Evaluation and Design for Wharf Berth Improvements
By: Xavier C. Barrett1, PE, Satrajit Das2, PhD, PE, M.ASCE, Richard C. Wells3, PE,
F. ASCE and Dennis K. Hoyle4, PE
1
Kleinfelder Southeast, Inc. Greensboro, NC, [email protected]; 2URS Corporation,
Raleigh, NC, [email protected]; 3Kleinfelder Southeast, Inc. Greensboro, NC,
[email protected]; 4URS Corporation, Raleigh, NC, [email protected]
ABSTRACT
The North Carolina State Ports Authority (NCSPA) has undertaken improvements
to its Morehead Port berth facilities, to better serve the deeper draft vessels utilized in
today’s shipping industry. In general, the improvements include a layer of tieback to
aid in carrying additional live loads from a recently constructed warehouse, and a
combination of additional layers of tiebacks and a king pile wall system to stabilize
the existing wall during the future removal of an existing rock berm and dredging
along approximately 213.4 meters (700 feet) of the mudline to a deeper elevation.
The Ports Authority commissioned an engineering study to determine a feasible
approach to accomplish the improvements.
The engineering study included a subsurface exploration to assess the subsurface
profile, engineering properties of the in-situ soils, and configuration of an existing
relieving platform behind the current wall system. The subsurface profile model was
utilized in a detailed soil-structure interaction (SSI) analysis to design the new wall
system. The design of the new wall system implemented innovative methodologies
and construction sequencing to minimize the effects of the new wall on the existing
wall system and warehouse.
This paper highlights key features of the project. It describes the background
information; project description; the design methodologies; soil exploration and
determination of engineering parameters used in the SSI analysis of the new wall
system; and incorporation of practical construction issues in the design.
BACKGROUND INFORMATION
The Morehead Port facilities are located at the juncture of Bogue Sound and the
Newport River in Carteret County, North Carolina. The project site is located within
the Coastal Plain physiographic province, which consists of Cenozoic-era deposits
overlying Mezozoic-era crystalline rocks and tilted sedimentary rocks in buried
Triassic-Jurassic basins.
The elevation of the berth deck averages 3.0 meters (10 feet) above the mean low
water level (MLWL). The berths have a current design dredge elevation of -10.0
meters (-33 feet) MLWL, with a rip rap protective dike at the mudline. The riprap
berm is at an elevation of 5.5 meters (-18 feet) MLWL. An aerial photograph of the
overall site is provided in Figure 1 below:
Berth 9
Berth 8
N
Project Limits
Figure 1. Overview of Ports Facility
Berths 8 and 9 are to be improved, for a distance of about 213.4 meters (700 feet).
The improvements are to consist of strengthening the existing sheet piles to carry an
additional live load surcharge due to a newly constructed warehouse, and increasing
the design dredge elevation to -12.2 meters (-40 feet) MLWL. The following steps
were implemented in the design process:
1. Review design and construction documents for original installation.
2. Evaluate any existing geotechnical information in the area.
3. Determine critical geotechnical aspects for suitability relative to short term and
long term performance.
4. Collect additional geotechnical information including soil test borings and/or
dilatometer soundings for in-situ strength and deformation analysis.
5. Perform detailed SSI analysis to determine the optimum location and size of the
tiebacks and/or the king pile system.
6. Perform slope stability analysis to ensure overall stability of the wharf structure.
PROJECT DESCRIPTION
One of the major constraints consisted of the presence of a recently constructed
warehouse approximately 21.3 meters (70 feet) east of Berths 8 and 9 that must
remain in operation during the improvements. The warehouse, when full, comprises
a significant load, in both magnitude and extent. The live load stored in the
warehouse is approximately 71.8 kPa (1,500 psf) over an area of about 234.6 meters
by 70.3 meters (770 by 230 feet). The pavement surface outside the warehouse is
also used for storage, with an expected loading of 38.3 kPa (800 psf).
Due to the configuration of the existing berth wall system and the specific
requirements of the clients, the improvements will be constructed in phases as shown
in Figure 2.
Phase 1 (Reduce warehouse live load surcharge effects on the existing sheet piles):
Install new wales and tieback anchors at an elevation of -1.5 meters (-5.0 feet)
designed to carry a force of 289 kN (65 kips) each, connecting to the existing sheet
pile wall as shown in Figure 3.
Phase 2 (Increase dredge depth):
Based on a feasibility study pertaining to increasing the dredge depth at Berths 8
and 9, the following options are under consideration to keep the stresses in the
existing sheet piles within allowable limits. The existing riprap provides the
necessary passive reaction to ensure overall stability and limit stresses in the sheet
piles. Therefore, the riprap will not be removed from the toe of the sheet pile at any
time during construction and dredging operation.
Two options are under
consideration for accomplishing Phase 2, as outlined below.
Option 1: Readjustment of the force in the tiebacks installed during Phase 1 to 467
kN (105 kips) each. The net anticipated maximum deflection in the sheet piles using
this option is expected to be more than 83 mm (3.3 inches). In order to reduce the
passive pressure build-up, this method is expected to be performed in at least two
stages during the dredging operation.
Option 2: Installation of a second row of tiebacks at an elevation of -3.4 meter (11.0 feet) designed to carry a force of 334 kN (75 kips) each, and re-adjustment of the
force in the tiebacks installed during Phase 1 to 147 kN (33 kips) each. The net
anticipated maximum deflection in the sheet piles using this option is expected to be
less than 53 mm (1.8 inches). In order to reduce the passive pressure build-up, this
method is expected to be performed in at least three stages during the dredging
operation.
Figure 2. Construction Sequence in Different Phases of the Project
Based on the actual field conditions and the contractor’s preferred means and
methods of construction, Option 3 is under consideration for situations where the
overall stability of the system may be compromised due to unintended disturbance or
removal of the rip rap from the toe of the sheet piles and/or removal of the battered
pile or other related factors. A detailed SSI analysis in conjunction with slope
stability analysis will dictate the preferred combination of construction options
suitable for Phase 2.
Option 3: Will consist of several stages, as outlined below.
Stage 1: Remove the existing pavement and crane rails. Excavate down to the
existing cap beam and remove the existing relieving platform.
Stage 2: Demolish the existing cap beam. Demolish the existing batter piles down
to the top of the existing rip rap berm. Remove the rip rap berm, stockpile the
materials for reuse.
Stage 3: Construct a new king pile combination wall in front of the existing sheet
pile wall. Connect new tieback anchors to the new combination wall.
Stage 4: Dredge to lower the mudline to the new elevation of -12.2 meters (-40
feet) MLWL. Reconstruct the rip rap berm with the previously stockpiled
materials.
Stage 5: Construct a new cap beam. Backfill and reconstruct the relieving
platform. Reinstall the crane rails and pavement.
DESIGN METHODOLOGY
Preliminary investigations indicated that the additional warehouse live load
surcharge would likely overload the existing sheet piles. The current improvements at
Berths 8 and 9 are directed towards designing an alternative means of lateral load
transfer with an objective to reduce the excessive loads on the existing sheet piles.
Based on a study of different alternatives, it was decided that tiebacks (pre-stressed
soil anchors) will be designed to reduce the loads on the sheet piles.
The need for a geotechnical investigation to determine the in-situ properties of the
existing coastal plain soils in the anticipated zone of tieback construction was
identified. The purpose of this study was to provide geotechnical parameters to be
utilized in the subsequent SSI analysis and design of the sheet pile wall, the tieback
anchors and the laterally loaded king piles. To accomplish this, four (4) Dilatometer
(DMT) soundings were performed within the study area to depths of 12.2 meters (40
feet) below the existing pavement surface, to assess the in-situ properties in the area
of the tieback anchors. In addition, the results of eight (8) soil test borings from
previous explorations were reviewed for correlations. The soil test borings had been
extended to depths of between 6.1 meters (20 feet) and 30.5 meters (100 feet) below
the existing pavement surface. Relatively undisturbed, Shelby Tube samples were
obtained from the coastal plain soils in selected borings and returned to the
laboratory.
The laboratory testing program included routine index property testing of selected
split spoon samples as well as testing of the Shelby tube samples. The index testing
consisted of moisture contents, Atterberg limits, and percent fines. The Shelby tube
samples were subjected to CU triaxial shear tests with pore pressure measurements to
determine the shear strength properties.
Following the subsurface investigations, a SSI analysis was carried out on a 2-D
finite element model using the general purpose structural analysis software package
GTStrudl. The analytical model was set-up iteratively to account for the non-linear
behavior of the coastal plain soil. The model accounted for the stiffness of the
structural components including the sheet pile, the battered pile, the tie back, and
numerically substituted the passive resistance of soil by estimated values of modulus
of sub-grade reaction. The model was subjected to anticipated active pressure acting
on the sheet pile. The presence of the relieving platform was considered in the stress
analysis. The screening effects of the 18-inch square piles located inboard of the
sheet piles were not considered in the computations.
SOIL EXPLORATION AND ENGINEERING PARAMETERS USED IN SSI
ANALYSIS
The ground profile in the study area consists of an upper zone of earth fill underlain
by a coastal plain soil profile, with groundwater at elevation 0.0 meters MLWL. The
earth fill is present to an elevation of about -10.7 meters (-35 feet) MLWL. The earth
fill primarily consists of sands with some intermittent clay layers. The sands ranged
from fine-grained sands to coarse-grained sands. The sands ranged from relatively
clean sands (SP) with little to no fines (silt and clay) content, to silty and clayey sands
(SM and SC) with appreciable fines content. Most of the sands contained a trace to
appreciable amounts of shell fragments, with some samples containing gravel and
rounded pebbles. Standard Penetration Test (SPT) N-values ranging from 0 to 23
blows per 0.3 meter (1 foot) were obtained in the fill, with the majority of values less
than 10. Liquid Limit values of the samples obtained in the fill ranged from 17 to 43
percent. Plasticity Index values of the samples obtained in the fill ranged from nonplastic to 26 percent. Undrained shear strength values obtained in the clayey fill
zones ranged from 21.7 to 63.4 kPa (0.45 ksf to 1.32 ksf). Dilatometer modulus (ED)
values obtained in the fill ranged from 729.2 kPa to 151310.7 kPa (15 to 3160 ksf),
with the majority of values less than about 17000 kPa (355 ksf). A statistical
summary of the dilatometer modulus values for the fill is provided below.
Table 1. Summary of Dilatometer ED Values in Fill
Low Value
High Value
Average
Std. Deviation
Median
729.2 kPa
151310.7 kPa 43568.2 kPa
37473.0 kPa
35002.0 kPa
(15.2 ksf)
(3160.2 ksf)
(909.9 ksf)
(782.6 ksf)
(731.0 ksf)
Below the earth fill, the majority of the coastal plain soils consist of clean sands
(SP) and silty sands (SM) with some layers of clayey sands (SC). The clean sands
(SP) and silty sands (SM) were found to be generally non-plastic in nature as tested in
the laboratory, while the clayey sands (SC) exhibited some moderate to high
plasticity with laboratory determined plasticity indices (PI) ranging from 10 to 51.
The clayey sands (SC) were typically encountered at elevations from -20.4 meters (67 feet) MLWL to -22.0 meters (-72 feet) MLWL and typically extended to the
boring termination depths. Standard Penetration Test (SPT) N-values ranging from 0
to 95 blows per 0.3 meter (1 foot) were obtained in the fill, with the majority of
values less than 16. Liquid limit values of the samples obtained in the coastal plain
soils ranged from 17 to 77 percent. Plasticity index values of the samples obtained in
the coastal plain soils ranged from non-plastic to 51 percent. Undrained shear
strength values obtained in the clayey Coastal Plain soils ranged from 11.4 kPa to
34.7 kPa (0.24 ksf to 0.72 ksf). Dilatometer modulus (ED) values obtained in the
Coastal Plain soil zones ranged from 1093.8 kPa to 86046.6 kPa (23 ksf to 1797 ksf),
with the majority of values less than 30000 kPa (627 ksf). A statistical summary of
the dilatometer modulus values for the fill is provided below.
Table 2. Summary of Dilatometer ED Values in Coastal Plain Soils
Low Value
High Value
Average
Std. Deviation
Median
1093.8 kPa
86046.6 kPa
27926.4 kPa
21128.2 kPa
25340.0 kPa
(22.8 ksf)
(1797.1 ksf)
(583.3 ksf)
(441.3 ksf)
(529.2 ksf)
The results of the triaxial shear tests on the Coastal Plain soil samples indicated
total and effective shear strength parameters as follows:
Table 3. Summary of Triaxial Shear Test Results
Sample Elevation
USCS
φ (deg) C (kPa) φ’ (deg)
-12.3 m (-40.4 ft)
SM
23
28.73
34
-13.7 m(-45.0 ft)
SM
19
38.30
38
-16.8 m (-55.0 ft)
SM
16
35.91
37
-15.2 m(-50.0 ft)
SM
17
28.73
36
-23.0 m (-75.5 ft)
SC
20
28.73
50
C’ (kPa)
0.0
0.0
0.0
0.0
0.0
The results of the field and laboratory testing were evaluated to determine the
proper parameters to be used in the design of the new tieback and the king pile
system. Engineering parameters were provided for the existing fill and the
underlying Coastal Plain soils. The parameter values provided included unit weight
(γ), effective friction angle (φ’), undrained shear strength (Su), vertical drained
constrained modulus (MDMT), modulus of subgrade reaction (Kh), and grout-to-soil
bond strength.
The unit weight values indicated by the dilatometer soundings did not appear to
correlate well with the SPT N-values (corrected for effective overburden pressure).
They appeared to better correlate with the unit weights from the undisturbed samples.
Therefore, the calculated values were considered to be nominal values for design.
The effective friction angle values indicated by the dilatometer soundings appeared
to correlate well with the results of the triaxial shear tests. Therefore, the calculated
values were considered to be upper bound values for design.
The undrained shear strength values indicated by the dilatometer soundings for the
clayey zones in the fill and Coastal Plain appeared to moderately correlate with the
SPT N-values. Therefore, the calculated values were considered to be upper bound
values for design.
The Kh values were determined using the method proposed by Marchetti. This
method consistently produced values greater than the method outlined by Bowles
(Bowles 1997) and thus provided a logical upper limit of the modulus of subgrade
reaction for the local soil conditions. However, in order to estimate the maximum
possible deflection and bending moment in the existing sheet pile, the values of Kh
used in the SSI analysis were based on the Bowle’s method.
The calculations to determine the vertical drained constrained modulus values were
adjusted based on the visual-manual classification of the split spoon samples removed
from the soil test borings.
A summary of the average design parameters obtained from the geotechnical analysis
is outlined below.
Table 4. Summary of Average Engineering Design Parameters
Depth (m)
γ (kg/m3)
φ’
Su (kPa) kh (kg/cm3) MDMT (kPa)
0.0
0.0 – 3.0
2000.0
36
19.5
291,360
0.0
3.0 – 9.1
1790.0
34
4.4
65,950
0.0
9.1 – 12.2
1810.0
30
2.7
40,250
0.0
12.2 – 15.2
1680.0
30
2.7
40,250
0.0
>15.2 (active)
1600.0
26
4.8
70,000
0.0
>15.2 (passive)
1600.0
26
4.8
70,000
The grout to soil bond strengths were correlated to the in-situ unit weight of the fill
soils. The design values are to be confirmed during the field testing program.
Table 5. Summary of Average Grout to Soil Bond Strength Parameters
Ultimate Bond Stress (kPa )
γ (kg/m3)
1441.7 – 1601.9 (90-100 pcf)
82.7 (1.73 ksf)
1617.9 – 1762.0 (101-110 pcf)
131.0 (2.74 ksf)
1778.1 – 1922.2 (111-120 pcf)
199.9 (4.17 ksf)
1938.2 – 2082.4 (121-130 pcf)
275.8 (5.76 ksf)
2098.4 – 2242.6 (131-140 pcf)
344.7 (344.73 ksf)
Figure 3. Tieback Installation Details in Phase I Construction
PRACTICAL CONSTRUCTION ISSUES
Due to the wide variation in the in-situ soil strengths and the presence of clayey soil
layers, field testing will be performed to confirm the design parameters. The field
testing will confirm the calculated parameters utilized in the design, and the results
can be utilized to remove excess conservatism from the design. Tieback locations
that result in the anchor zones in the intermittent clayey layers should be tested to
assess the effects of creep.
The tiebacks will be designed for a marine environment. The tieback location with
respect to the MLWL will require drilling and installation methods that are suitable
for underwater applications. The anchors will require double corrosion protection
with factory pre-grouted encapsulation of the bar within a corrugated plastic sheath.
Protection of the anchor head shall be provided by a fully enclosed assembly. Based
on project requirements including future needs of lowering the mud-line, the anchor
system shall be re-stressable. The free length of the anchor must not be bonded to the
surrounding grout. The load in the bar tendons shall be adjusted by turning the
anchor nut. Tieback details are included in Figure 3.
The estimated cost of construction for installation of the wales and the tiebacks in
Phase 1 is approximately US$1,980 per linear meter (US$6,500 per linear foot) and
the additional cost for installation of the king pile system and the lowering of the
dredge depth in Phase 2 is more than US$3,200 per linear meter (US$10,500 per
linear foot).
REFERENCES
Bowles, Joseph E. (1997) Foundation Analysis and Design, Fifth Edition.
Ebeling, Azene, and Strom. (2002) Simplified Procedures for the Design of Tall,
Flexible Anchored Tieback Walls. U.S. Army Corps of Engineers. ERDC/ITL TR02-09.
FHWA (1999) Ground Anchors and Anchored Systems. FHWA Geotechnical
Engineering Circular No. 4. FHWA Report No. FHWA-IF-99-015.
FHWA (2002) Evaluation of Soil and Rock Properties. FHWA Geotechnical
Engineering Circular No. 5. FHWA Report No. FHWA-IF-02-034.
Pelnick, et al. (1999) Foundation Design Applications of CPTU and DMT Tests in
Atlantic Coastal Plain Virginia. Washington, D.C. Transportation Research Board,
78th Annual Meeting, January 10-14, 1999: Paper No. 990794.
ISSMGE Technical Committee 16. “The Flat Dilatometer test (DMT) in Soil
Investigations.” Proceedings of IN SITU 2001, International Conference on In situ
Measurement of Soil Properties, Bali, Indonesia